Fire and Steel Construction.
The Steel Construction Institute.
3 THE CARDINGTON FIRE TESTS
3.1 Research programme
In September 1996, a programme of fire tests was completed in the UK at the Building Research Establishment's Cardington Laboratory. The tests were carried out on an eight-storey composite steel-framed building that had been designed and constructed as a typical multi-storey office building. The purpose of the tests was to investigate the behaviour of a real structure under real fire conditions and to collect data that would allow computer programs for the analysis of structures in fire to be verified.
The test building (see Figure A.1.1) was designed to be a typical example of both the type of braced structure and the load levels that are commonly found in the UK. In plan, the building covered an area of 21 m x 45 m and had an overall height of 33 m. The beams were designed as simply supported acting compositely with a 130 mm floor slab. Normally a building of this type would be required to have 90 minutes fire resistance. Fin-plates were used for the beam-to-beam connections and flexible end plates for the beam-to-column connections. The structure was loaded using sandbags distributed over each floor to simulate typical office loading.
There were two projects in the research programme. One project was funded by Corus (formerly British Steel) and the European Coal and Steel Community (ECSC), and the other was funded by the UK Government via the Building Research Establishment (BRE). Other organisations involved in the research programme included Sheffield University, TNO (The Netherlands), CTICM (France) and The Steel Construction Institute. Fire tests took place between January 1995 and July 1996. The tests were carried out on various floors; the location of each test is shown on the floor plan in Figure B.3.1.
Figure B.3.1 Test locations
More details of the test building, including member sizes, are given in Appendix 3.
Test 1 involved a single secondary beam and the surrounding floor slab which was heated by a purpose-built gas-fired furnace. Test 2 was also heated using gas, and was conducted on a plane frame spanning across the building on one floor; the test included primary beams and associated columns. Tests 3, 4 and 5 involved compartments of various sizes subjected, in each case, to a natural fire fuelled by timber cribs. The columns in these tests were protected up to the underside of the floor slab and the beams and floor slab were left unprotected. The last, and most severe, test was a demonstration using furniture and contents typically found in modern offices.
A detailed description of the tests has been published . The complete test data, in electronic form with accompanying instrument location maps, is available for Tests 1, 2, 3 and 6 from Corus RD&T (Swinden Technology Centre) and for Tests 4 and 5 from BRE [14,15].
3.2 Test 1: Restrained beam
The test was carried out on the seventh floor of the building. A purpose-built gas fired furnace, 8.0 m long and 3.0 m wide, was designed and constructed to heat a secondary beam (D2/E2) spanning between two columns and part of the surrounding structure. The beam was heated over the middle 8.0 m of its 9.0 m length, thus keeping the connections relatively cool. The purpose of the test was to investigate the behaviour of a heated beam surrounded by an unheated floor slab and study the restraining effect of the unheated parts of the structure.
The beam was heated at between 3 and 10°C per minute until temperatures approaching 900°C were recorded. At the peak temperature, 875�C in the lower flange, the mid span deflection was 232 mm (span/39) (see Figure B.3.2). On cooling, the mid-span deflection recovered to 113 mm.
Figure B.3.2 Central displacement and maximum temperature in restrained beam test
The contrast between the behaviour of this beam and a similar unprotected beam tested in the standard fire test under a similar load  is shown in Figure B.3.3. The `runaway' displacement typical of simply supported beams in the standard test did not occur to the beam in the building frame even though, at a temperature of about 900�C, structural steel retains only about 6% of its yield strength at ambient temperature. This gives an indication that standard fire test results are not a reliable measure of real building performance in fire.
Figure B.3.3 Central displacement and maximum temperature in standard fire test and restrained beam test
During the test, local buckling occurred at both ends of the test beam, just inside the furnace wall (see Figure B.3.4).
Figure B.3.4 Flange buckling in restrained beam
Visual inspection of the beam after the test showed that the end-plate connection at both ends of the beam had fractured near, but outside, the heat-affected zone of the weld on one side of the beam. This was caused by thermal contraction of the beam during cooling, which generated very high tensile forces. Although the plate sheared down one side, this mechanism relieved the induced tensile strains, with the plate on the other side of the beam retaining its integrity and thus providing shear capacity to the beam. The fracture of the plate can be identified from the strain gauge readings, which indicate that during cooling the crack progressed over a period of time rather than by a sudden fracture.
3.3 Test 2: Plane frame
This test was carried out on a plane frame consisting of four columns and three primary beams spanning across the width of the building (B1/B4).
A gas-fired furnace 21 m long x 2.5 m wide x 4.0 m high was constructed using blockwork across the full width of the building.
The primary and secondary beams, together with the underside of the composite floor, were left unprotected. The columns were fire protected to a height at which a suspended ceiling might be installed (although no such ceiling was present). This resulted in the top 800 mm of the columns, which incorporated the connections, being unprotected.
The rate of vertical displacement at midspan of the 9 m span steel beam increased rapidly between approximately 110 and 125 minutes (see Figure B.3.5). This was caused by vertical displacements of its supporting columns. The exposed areas of the internal columns squashed by approximately 180 mm (see Figure B.3.6). The temperature of the exposed part of the column was approximately 670°C when squashing commenced.
Figure B.3.5 Maximum vertical displacement of central 9 m beam and temperature of exposed top section of internal column
The squashing of the column resulted in all the floors above the fire compartment being displaced downwards by approximately 180 mm. To avoid this behaviour, columns in later tests were protected over their full height.
Figure B.3.6 Squashed column head following the test
On both sides of the primary beams, the secondary beams were each heated over a length of approximately 1.0 m. After the test, investigation showed that many of the bolts in the fin-plate connections had sheared (see Figure B.3.7). The bolts had only sheared on one side of the primary beam. In a similar manner to the fracturing of the plate in Test 1, the bolts sheared due to thermal contraction of the beam during cooling. The thermal contraction generated very high tensile forces, which were relieved once the bolts sheared in the fin-plate on one side of the primary beam.
Figure B.3.7 Fin-plate connection following test
3.4 Test 3: Corner
The objective of this test was to investigate the behaviour of a complete floor system and, in particular, the role of `bridging' or membrane action of the floor in providing alternative load paths as the supporting beams lose strength. Using concrete blockwork, a compartment 10 m wide x 7.6 m deep was constructed in one corner of the first floor of the building (E2/F1).
To ensure that the compartment walls did not contribute to supporting the applied loads, all the restraints and ties in the gable wall and the top layer of blockwork were removed. The mineral fibre board in the expansion joints was replaced with a ceramic blanket.
The wind posts were detached from the edge beam above the opening.
All columns, beam-to-column connections and edge beams were fire protected.
The fire load was 45 kg/m2, in the form of timber cribs. This fire load is quite high and is equivalent to the 95% fractile for office buildings. Fire safety engineering calculations are normally based on the 80% fractile. Ventilation was provided by a single 6.6 m wide x 1.8 m high opening. The peak atmosphere temperature recorded in the compartment was 1071�C.
The maximum steel temperature was 1014�C, in the inner beam E2/F2. The maximum vertical displacement of 428 mm (just less than span/20) occurred at the centre of the secondary beam, which had a peak temperature of 954�C. On cooling this recovered to a permanent displacement of 296 mm. The variations of deflection and temperature with time are shown in Figure B.3.8.
Figure B.3.8 Maximum vertical displacement and temperature of secondary beam
All the combustible material within the compartment was consumed by the fire. The structure behaved extremely well with no signs of collapse (see Figure B.3.9).
Buckling occurred in the proximity of some of the beam-to-column connections but unlike Test 2, bolts in the connections did not suffer shear failure. This might indicate either that the high tensile forces did not develop or that the connection had adequate ductility to cope with the tensile displacements.
Figure B.3.9 View of structure following test
3.5 Test 4: Corner
This test was carried out on the second floor, in a corner bay (E4/F3) with an area of 54 m2. The internal boundaries of the compartment on gridlines E and 3 were constructed using steel stud partitions with fire resistant board. The stud partition was specified to have 120 minutes fire resistance, with a deflection head of 15 mm. An existing full-height blockwork wall formed the boundary on the gable wall on gridline F; the outer wall, gridline 4, was glazed above 1 metre of blockwork. The compartment was totally enclosed with all windows and doors closed. The columns were fire protected up to the underside of the floor slab, including the connections but, unlike Test 3, the lintel beam (E4/F4) was unprotected and the wind posts above it remained connected. Twelve timber cribs were used to give a fire load of 40 kg/m2.
The development of the fire was largely influenced by the lack of oxygen within the compartment. After an initial rise in temperature, the fire died down and continued to smoulder until, after 55 minutes, the fire brigade intervened to vent the compartment by removal of a single pane of glazing. This resulted in a small increase in temperature followed by a decrease. A second pane, immediately above the first, was broken at 64 minutes and temperatures began to rise steadily; between 94 and 100 minutes the remaining panes shattered. This initiated a sharp increase in temperature that continued as the fire developed. The maximum recorded atmosphere temperature in the centre of the compartment was 1051°C after 102 minutes (see Figure B.3.10). The maximum steel temperature of 903°C was recorded after 114 minutes in the bottom flange of the central secondary beam.
The maximum slab displacement was 269 mm and occurred in the centre of the compartment after 130 minutes. This recovered to 160 mm after the fire.
The unprotected lintel beam, E4/F4, was observed during the test to be completely engulfed in fire. However, the maximum temperature of this beam was only 680°C, with a corresponding maximum displacement of 52 mm, recorded after 114 minutes. This small displacement was attributed to the additional support provided by wind posts above the compartment, which acted in tension during the test. The tops of these wind posts were fixed by bolts through slotted holes 80 mm long. Thus, a large proportion of the 52 mm displacement could be attributed to the bolt slippage. Unfortunately, the position of the bolts was not recorded before the test.
The internal compartment walls were constructed directly under unprotected beams and performed well. Their integrity was maintained for the duration of the test. On removal of the wall, it could be seen that one of the beams had distortionally buckled over most of its length. This was caused by the high thermal gradient through the cross section of the beam (caused by the positioning of the compartment wall), together with high restraint to thermal expansion.
No local buckling occurred in any of the beams, and the connections showed none of the characteristic signs of high tensile forces that were seen on cooling in the other tests.
Figure B.3.10 Furnace gas temperatures recorded in Test 4
Figure B.3.11 Maximum flange temperature of internal beam and edge beam
3.6 Test 5: Large compartment
This test was carried out between the second and third floor, with the fire compartment extending over the full width of the building, covering an area of 340 m2.
The fire load of 40 kg/m2 was provided by timber cribs arranged uniformly over the floor area. The compartment was constructed by erecting a fire resistant stud and plasterboard wall across the full width of the building and by constructing additional protection to the lift shaft. Double glazing was installed on two sides of the building, but the middle third of the glazing on both sides of the building was left open. All the steel beams, including the edge beams, were left unprotected. The internal and external columns were protected up to and including the connections.
The ventilation condition governed the severity of the fire. There was an initial rapid rise in temperature as the glazing was destroyed, creating large openings on both sides of the building. The large ventilation area in two opposite sides of the compartment gave rise to a fire of long duration but lower than expected temperatures. The maximum recorded atmosphere temperature was 746°C, with a maximum steel temperature of 691°C, recorded at the centre of the compartment. The structure towards the end of the fire is shown in Figure B.3.12.
The maximum slab displacement reached a value of 557 mm. This recovered to 481 mm when the structure cooled.
Extensive local buckling occurred in the proximity of the beam-to-beam connections. On cooling, a number of the end-plate connections fractured down one side. In one instance the web detached itself from the end-plate so that the steel-to-steel connection had no shear capacity. This caused large cracks within the composite floor above this connection, but no collapse occurred, with the beam shear being carried by the composite floor slab.
Figure B.3.12 Deformed structure during fire
Figure B.3.13 Maximum and average recorded atmosphere temperature
3.7 Test 6: The office demonstration test
The aim of this test was to demonstrate structural behaviour in a realistic fire scenario.
A compartment 18 m wide and up to 10 m deep with a floor area of 135 m2, was constructed using concrete blockwork. The compartment represented an open plan office and contained a series of work-stations consisting of modern day furnishings, computers and filing systems (see Figure B.3.14). The test conditions were set to create a very severe fire by incorporating additional wood/plastic cribs to create a total fire load of 46 kg/m2 (less than 5% of offices would exceed this level) and by restricting the window area to the minimum allowed by regulations for office buildings. The fire load was made up of 69% wood, 20% plastic and 11% paper. The total area of windows was 25.6 m2 (19% of the floor area) and the centre portion of each window, totalling 11.3 m2, was left unglazed to create the most pessimistic ventilation conditions at the start of the test.
Figure B.3.14 Office before test
Within the compartment, the columns and the beam-to-column connections were fire protected. Both the primary and secondary beams, including all the beam-to-beam connections, remained totally exposed.
The wind posts were left connected to the edge beams, and thus gave some support during the fire.
The maximum atmosphere temperature was 1213°C and the maximum average temperature was approximately 900°C (see Figure B.3.15). The maximum temperature of the unprotected steel was 1150°C. The maximum vertical displacement was 640 mm, which recovered to 540 mm on cooling (see Figure B.3.16). The peak temperature of the lintel beams, above the windows, was 813�C. All the combustible material in the compartment was completely burnt, including the contents of the filing cabinets. Towards the back of the compartment, the floor slab deflected and rested on the blockwork wall. The structure showed no signs of failure.
An external view of the fire near its peak is shown in Figure B.3.17. The structure following the fire is shown in Figure B.3.18 and Figure B.3.19. Figure B.3.18 shows a general view of the burned out compartment and Figure B.3.19 shows the head of one of the columns. During the test, the floor slab cracked around one of the column heads (see Figure B.3.20). This behaviour is discussed in Section 5.1.
Figure B.3.15 Measured atmosphere temperature
Figure B.3.16 Maximum steel temperature and vertical displacement
Figure B.3.17 External view of fire
Figure B.3.18 Compartment following fire
Figure B.3.19 Column head showing buckled beams
Figure B.3.20 Cracked floor slab in region of non-overlapped mesh
3.8 General comments on observed behaviour in all tests
In all tests, the structure performed very well and overall structural stability was maintained.
The performance of the whole building in fire is manifestly very different from the behaviour of single unrestrained members in the standard fire test. It is clear that there are interactions and changes in load-carrying mechanisms in real structures that dominate the way they behave; it is entirely beyond the scope of the simple standard fire test to reproduce or assess such effects.
The Cardington tests demonstrated that modern steel frames acting compositely with steel deck floor slabs have a coherence that provides a resistance to fire far greater than that normally assumed. This confirms evidence from other sources.
4 EVIDENCE FROM OTHER FIRES AND OTHER COUNTRIES
In recent years, two building fires in England (Broadgate and Churchill Plaza) have provided the opportunity to observe how modern steel-framed buildings perform in fire. The experience from these fires was influential in stimulating thought about how buildings might be designed to resist fire and in bringing about the Cardington experiments.
Evidence of building behaviour is also available from large-scale fire tests in Australia and Germany. In both Australia and New Zealand, design approaches that allow the use of unprotected steel in multi-storey, steel-framed buildings have been developed.
In 1990, a fire occurred in a partly completed 14-storey office block on the Broadgate development in London . The fire began inside a large site hut on the first level of the building. Fire temperatures were estimated to have reached over 1000°C.
The floor was constructed using composite long-span lattice trusses and composite beams supporting a composite floor slab. The floor slab was designed to have 90 minutes fire resistance. At the time of the fire, the building was under construction and the passive fire protection to the steelwork was incomplete. The sprinkler system and other active measures were not yet operational.
After the fire, a metallurgical investigation concluded that the temperature of the unprotected steelwork was unlikely to have exceeded 600°C. A similar investigation on the bolts used in the steel-to-steel connections concluded that the maximum temperature reached in the bolts, either during manufacture or as a consequence of the fire, was 540°C.
The distorted steel beams had permanent deflections of between 270 mm and 82 mm. Beams with permanent displacements at the higher end of that range showed evidence of local buckling of the bottom flange and web near their supports. From this evidence, it was concluded that the behaviour of the beams was influenced strongly by restraint to thermal expansion. This restraint was provided by the surrounding structure, which was at a substantially lower temperature than the fire-affected steel. Axial forces were induced into the heated beams resulting in an increase in vertical displacement due to the P-delta effect. The buckling of the lower flange and web of the beam near its supports was due to a combination of the induced axial force and the negative moment caused by the fixity of the connection.
Although the investigation showed the visually unfavourable effects of restraint on steel beams, the possible beneficial effects were not evident because only relatively low steel temperatures were reached during the fire. The beneficial effects that could have developed were catenary action of the beams and bridging or membrane action of the composite slab.
The fabricated steel trusses spanned 13.5 m and had a maximum permanent vertical displacement of 552 mm; some truss elements showed signs of buckling. It was concluded that the restraint to thermal expansion provided by other elements of the truss, combined with non-uniform heating, caused additional compressive axial forces, which resulted in buckling.
At the time of the fire, not all the steel columns were fire protected. In cases where they were unprotected, the column had deformed and shortened by approximately 100 mm (see Figure B.4.1). These columns were adjacent to much heavier columns that showed no signs of permanent deformation. It was thought that this shortening was a result of restrained thermal expansion. The restraint to thermal expansion was provided by a rigid transfer beam at an upper level of the building, together with the columns outside the fire affected area.
Figure B.4.1 Buckled column and deformed beams at Broadgate
Although some of the columns deformed, the structure showed no signs of collapse. It was thought that the less-affected parts of the structure were able to carry the additional loads that were redistributed away from the weakened areas.
Following the fire, the composite floor suffered gross deformations with a maximum permanent vertical displacement of 600 mm (see Figure B.4.2). Some failure of the reinforcement was observed. In some areas, the steel profiled decking had debonded from the concrete. This was considered to be caused mainly by steam release from the concrete, together with the effects of thermal restraint and differential expansion.
A mixture of cleat and end-plate connections was used. Following the fire, none of the connections was observed to have failed although deformation was evident. In cleated connections, there was some deformation of bolt holes. In one end-plate connection, two of the bolts had fractured; in another, the plate had fractured down one side of the beam but the connection was still able to transfer shear. The main cause of deformation was thought to be due to the tensile forces induced during cooling.
Following the fire, structural elements covering an area of approximately 40 m x 20 m were replaced, but it is important to note that no structural failure had occurred and the integrity of the floor slab was maintained during the fire. The direct fire loss was in excess of �25M, of which less than �2M was attributed to the repair of the structural frame and floor damage; the other costs resulted from smoke damage. Structural repairs were completed in 30 days.
Figure B.4.2 View of deformed floor above the fire (the maximum deflection was about 600 mm)
4.2 Churchill Plaza Building, Basingstoke
In 1991, a fire took place in the Mercantile Credit Insurance Building, Churchill Plaza, Basingstoke. The 12-storey building was constructed in 1988. The columns had board fire protection and the composite floor beams had spray- applied protection. The underside of the composite floor was not fire protected. The structure was designed to have 90 minutes fire resistance.
The fire started on the eighth floor and spread rapidly to the ninth and then the tenth floor as the glazing failed. During the fire, the fire protection performed well and there was no permanent deformation of the steel frame. The fire was believed to be comparatively `cool' because the failed glazing allowed a cross wind to increase the ventilation. The protected connections showed no deformation.
In places, the dovetail steel decking showed some signs of debonding from the concrete floor slab. This had also been observed in the Broadgate fire. A load test was conducted on the most badly affected area. A load of 1.5 times the total design load was applied. The test showed that the slab had adequate load- carrying capacity and could be reused without repair.
The protected steelwork suffered no damage. The total cost of repair was in excess of �15M, most of which was due to smoke contamination, as in the Broadgate fire. Sprinklers were installed in the refurbished building.
Figure B.4.3 Churchill Plaza, Basingstoke following the fire
4.3 Australian fire tests
BHP, Australia's biggest steel maker, has been researching and reporting [18,19] fire-engineered solutions for steel-framed buildings for many years. A number of large-scale natural fire tests has been carried out in specially constructed facilities at Melbourne Laboratory, representing sports stadia, car parks and offices. The office test programme focussed on refurbishment projects that were to be carried out on major buildings in the commercial centre of Melbourne.
4.3.1 William Street fire tests and design approach
A 41-storey building in William Street in the centre of Melbourne was the tallest building in Australia when it was built in 1971. The building was square on plan, with a central square inner core. A light hazard sprinkler system was provided. The steelwork around the inner core and the perimeter steel columns were protected by concrete encasement. The beams and the soffit of the composite steel deck floors were protected with asbestos-based material. During a refurbishment programme in 1990, a decision was made to remove the hazardous asbestos.
The floor structure was designed to serviceability rather than strength requirements. This meant that there was a reserve of strength that would be very beneficial to the survival of the frame in fire, as higher temperatures could be sustained before the frame reached its limiting condition.
At the time of the refurbishment, the required fire resistance was 120 minutes. Normally this would have entailed the application of fire protection to the steel beams and to the soffit of the very lightly reinforced slab (Australian regulations have been revised and now allow the soffit of the slab to be left unprotected for 120 minutes fire resistance). In addition, the existing light hazard sprinkler system required upgrading to meet the prevailing regulations.
During 1990, the fire resistance of buildings was subject to national debate; the opportunity was therefore taken to conduct a risk assessment to assess whether fire protecting the steelwork and upgrading the sprinkler system was necessary for this building. Two assessments were made. The first was made on the basis that the building conformed to current regulations with no additional safety measures; the second was made assuming no protection to the beams and soffit of the slab, together with the retention of the existing sprinkler system. The effect of detection systems and building management systems were also included in the second assessment. The authorities agreed that if the results from the second risk assessment were at least as favourable as those from the first assessment, the use of the existing sprinkler system and unprotected steel beams and composite slabs would be considered acceptable.
A series of four fire tests was carried out to obtain data for the second risk assessment. The tests were to study matters such as the probable nature of the fire, the performance of the existing sprinkler system, the behaviour of the unprotected composite slab and castellated beams subjected to real fires, and the probable generation of smoke and toxic products.
The tests were conducted on a purpose-built test building at the Melbourne Laboratories of BHP Research (see Figure B.4.4). This simulated a typical storey height 12 m x 12 m corner bay of the building. The test building was furnished to resemble an office environment with a small, 4 m x 4 m, office constructed adjacent to the perimeter of the building. This office was enclosed by plasterboard, windows, a door, and the facade of the test building. Imposed loading was applied by water tanks.
Figure B.4.4 BHP test building and fire test
Four fire tests were conducted. The first two were concerned with testing the performance of the light hazard sprinkler system. In Test 1, a fire was started in the small office and the sprinklers were activated automatically. This office had a fire load of 52 kg/m2. The atmosphere temperatures reached 60°C before the sprinklers controlled and extinguished the fire. In Test 2, a fire was started in the open-plan area midway between four sprinklers. This area had a fire load of 53.5 kg/m2. The atmosphere temperature reached 118°C before the sprinklers controlled and extinguished the fire. These two tests showed that the existing light hazard sprinkler system was adequate.
The structural and thermal performance of the composite slab was assessed in Test 3. The supporting beams were partially protected. The fire was started in the open plan area and allowed to develop with the sprinklers switched off. The maximum atmosphere temperature reached 1254°C. The fire was extinguished once it was considered that the atmosphere temperatures had peaked. The slab supported the imposed load. The maximum temperature recorded on the top surface of the floor slab was 72°C. The underside of the slab had been partially protected by the ceiling system, which remained substantially in place during the fire.
In Test 4, the steel beams were left unprotected and the fire was started in the small office. The fire did not spread to the open-plan area despite manual breaking of windows to increase the ventilation. Therefore fires were ignited from an external source in the open-plan area. The maximum recorded atmosphere temperature was 1228°C, with a maximum steel beam temperature of 632°C above the suspended ceiling. The fire was extinguished when it was considered that the atmosphere temperatures had peaked. Again, the steel beams and floor were partially shielded by the ceiling. The central displacement of the castellated beam was 120 mm and most of this deflection was recovered when the structure cooled to ambient temperature.
Three unloaded columns were placed in the fire compartment to test the effect of simple radiation shields. One column was shielded with galvanized steel sheet, one with aluminised steel sheet and one was an unprotected reference column. The maximum recorded column temperatures were 580�C, 427�C and 1064�C respectively, suggesting that simple radiation shields might provide sufficient protection to steel members in low fire load conditions.
It was concluded from the four fire tests that the existing light hazard sprinkler system was adequate and that no fire protection was required to the steel beams or soffit of the composite slab. Any fire in the William Street building should not deform the slab or steel beams excessively, provided that the steel temperatures do not exceed those recorded in the tests.
The temperature rise in the steel beams was affected by the suspended ceiling system, which remained largely intact during the tests.
The major city centre office building that was the subject of the technical investigation was owned by Australia's largest insurance company, which had initiated and funded the test programme. It was approved by the local authority without passive fire protection to the beams but with a light hazard sprinkler system of improved reliability and the suspended ceiling system that had proved to be successful during the test programme.
4.3.2 Collins Street fire tests
This test rig was constructed to simulate a section of a proposed steel-framed multi-storey building in Collins Street, Melbourne. The purpose of the test was to record temperature data in fire resulting from combustion of furniture in a typical office compartment.
The compartment was 8.4 m x 3.6 m and filled with typical office furniture, which gave a fire load between 44 and 49 kg/m2. A non-fire-rated suspended ceiling system was installed, with tiles consisting of plaster with a fibreglass backing blanket. An unloaded concrete slab formed the top of the compartment. During the test, temperatures were recorded in the steel beams between the concrete slab and the suspended ceiling. The temperatures of three internal free-standing columns were also recorded. Two of these columns were protected with aluminium foil and steel sheeting, acting simply as a radiation shield; the third remained unprotected. Three unloaded external columns were also constructed and placed 300 mm from the windows around the perimeter of the compartment.
The non-fire-rated ceiling system provided an effective fire barrier, causing the temperature of the steel beams to remain low. During the test the majority of the suspended ceiling remained in place. Atmosphere temperatures below the ceiling ranged from 831°C to 1163°C, with the lower value occurring near the broken windows. Above the ceiling, the air temperatures ranged from 344°C to 724°C, with higher temperatures occurring where the ceiling was breached. The maximum steel beam temperature was 470°C.
The unloaded indicative internal columns reached a peak temperature of 740°C for the unprotected case and below 403°C for the shielded cases. The bare external columns recorded a peak temperature of 490°C.
This fire test showed that the temperatures of the beams and external columns were sufficiently low to justify the use of unprotected steel and, as in the William Street tests, the protection afforded by a non-fire-rated suspended ceiling was beneficial. The role of ceilings is discussed further in Section 5.7.3.
4.3.3 Conclusions from Australian research
The Australian tests and associated risk assessments concluded that, provided that high-rise office buildings incorporate a sprinkler system with a sufficient level of reliability, the use of unprotected beams would offer a higher level of life safety than similar buildings that satisfied the requirements of the Building Code of Australia by passive protection. Up to the beginning of 1999, six such buildings between 12 and 41 storeys were approved in Australia.
4.4 German fire test
In 1985, a fire test was conducted on a four storey steel-framed demonstration building constructed at the Stuttgart-Vaihingen University in Germany . Following the fire test, the building was used as an office and laboratory.
The building was constructed using many different forms of steel and concrete composite elements. These included water filled columns, partially encased columns, concrete filled columns, composite beams and various types of composite floor.
The main fire test was conducted on the third floor, in a compartment covering approximately one-third of the building. Wooden cribs provided the fire load and oil drums filled with water provided the gravity load. During the test, the atmosphere temperature exceeded 1000°C, with the floor beams reaching temperatures up to 650°C. Following the test, investigation of the beams showed that the concrete-infilled webs had spalled in some areas exposing the reinforcement. However, the beams behaved extremely well during the test with no significant permanent deformations following the fire. The external columns and those around the central core showed no signs of permanent deformation. The composite floor reached a maximum displacement of 60 mm during the fire and retained its overall integrity.
Following the fire, the building was reinstated. The refurbishment work involved the complete replacement of the fire damaged external wall panels, the damaged portions of steel decking to the concrete floor slab, and the concrete infill to the beams. Overall, it was shown that refurbishment to the structure was economically possible.
4.5 New Zealand design approach
The New Zealand Heavy Engineering Research Association (HERA) has published a draft guide on the design of multi-storey buildings . The guide is more ambitious in its scope than the guidance presented in Part A of this publication but is more conservative in some areas. In particular, it requires the ends of primary beams to be fire protected. (This advice is not included in the Part A recommendations because, although the bottom flange of some beams buckled at Cardington, no collapse occurred, and because finite element modelling by many researchers has shown that preserving the bending continuity of beams is not vital.)
The HERA guide details a method of design that allows the capacity of an unprotected composite floor system, such as that used in the Cardington tests, to be checked. The design method allows each individual element to be checked to ensure that it is capable of transferring the loads to which it is likely to be subjected at elevated temperatures. Useful guidance on the detailing of elements in order to allow the composite action to be fully developed is also given.
4.5.1 Summary of HERA design method
The design method is based on the prediction of atmospheric temperatures based on a realistic pattern of fire growth and spread. The method uses the parametric temperature-time curves that are modified versions of those given in the Eurocode, ENV 1991-2-2 . The time-temperature relationship for a fully developed fire can be calculated, given the dimensions of the compartment, the thermal conductivity of the linings, the area of openings and the fire load energy density. These relationships are applied to design areas within which the fire conditions are considered to exist. Defined rules allow the spread of fire around the compartment to be modelled.
Methods of accounting for the beneficial effects of ceilings in shielding the unprotected steelwork from fire are presented, based on the performance of the BHP William Street tests. The design guidance is based on indicative times for which the suspended ceiling would remain in place under fully developed fire conditions.
The temperature-time relationship of the steel is determined from the net heat flux from the fire to the surface of the steel member, including thermal convection and radiation based on ENV 1991-2-2. The temperature of the bottom flange of the steel beams is calculated first; temperatures of other steel components are calculated relative to these temperatures.
The load-carrying capacity of the slab and secondary beam system at elevated temperature is calculated using a combination of yield line analysis and tensile membrane action. The capacity given by the yield line analysis will usually not be sufficient to resist the design load in fire conditions, so the strength enhancement due to the development of tensile membrane action is utilised. The strength enhancement is a function of the deflection in the slab. As the required strength enhancement is known, this is used to determine the slab deflection at elevated temperature. Limits are applied to the level of strength enhancement obtained from membrane action and the maximum allowable deflection, in order to prevent undesirable failure modes such as fracture of reinforcement.
An important feature of the method is that the floor slab should be reinforced with ductile reinforcement. It is assumed that all reinforcement, including shrinkage and temperature mesh, fulfils an integral role in maintaining load- carrying capacity. The slab reinforcement must therefore be capable of a minimum of 10% elongation. A mesh of 6 mm diameter bars at 150 mm centres, giving an area of 188 mm2/m, is recommended. The mesh used is normally known in New Zealand as `seismic mesh'. Although its minimum elongation is 10% it is said to have an average elongation at failure of 20%. Reinforcement must be properly lapped.
Columns are required to be fully fire protected, as are the ends and connections of primary beams.
In fire conditions, the column is considered to be subject to normal gravity loading, plus some out-of-balance moments generated by different magnitudes of fire-induced rotation at the two primary connections together with some thermal induced axial force because of restraint. As the columns are fully fire- protected, their strength will remain close to their ambient temperature capacity and the thermal expansion will be low; the induced force due to restraint will therefore be small. No specific fire limit state design rules are given for columns. It is considered that columns that resist the ambient temperature conditions will be adequate under the reduced load of the fire limit state.
The main conclusion that can be drawn from the New Zealand approach is that a complete methodology has been developed that includes both fire and structural behaviour. The total approach has not been checked by the authors and cannot therefore be endorsed. Some elements of it are, as yet, unproven and parts of the New Zealand method (protecting the ends of beams) are considered unnecessary. There is no doubt, however, that the existence of a complete, verified, procedure of this type would be a great benefit and will be one of the objectives of the next stage of development (Level 2) in the UK.
5 OBSERVATIONS AND COMMENTS
From the observed structural behaviour at Cardington, Broadgate and other large-scale natural fires, some general observations can be made. These observations and evidence from mathematical modelling have led to the recommendations in Part A.
Modelling the behaviour of the structure in the Cardington tests has been undertaken by several universities. This work is described in papers published by The University of Sheffield , The University of Edinburgh with Corus, Swinden Technology Centre and Imperial College , and BRE .
5.1 Floor slabs
Observations from Cardington and building fires such as Broadgate have shown that the composite slab plays a crucial role in fire. It fulfils its separating functions by preventing the spread of fire. It is often the main load-carrying element, and it assists greatly in the maintenance of structural integrity through in-plane diaphragm action. This was seen clearly in the Broadgate fire where, despite vertical displacements of about 600 mm, horizontal displacements were negligible.
The composite floors in the Cardington frame suffered extensive cracking as a result of the fire tests, however their separating function was generally maintained during the tests. In Test 6, large cracks occurred around the internal column (see Figure B.3.20), which created gaps in the floor slab. The evidence suggests that these cracks occurred during cooling, possibly when the steel connection below the slab partially failed. Investigation of the slab after the test showed that the reinforcement had not been lapped correctly, and was just `butted' together.
The slab resists vertical loads in two ways. For normal design, and also in the established methods of fire design, e.g. BS5950-8 , the slab is assumed to act as a one-way spanning beam, resisting loads through bending and shear. The slab normally spans between the beams. Evidence from fire resistance tests and from some actual fires indicates that the bending tension is carried more by the reinforcement than by the profiled steel deck. At Cardington and Broadgate, the beams were not fire protected. This had the effect of greatly increasing the distance the floor slab was spanning, but no collapse occurred. The reason for this is complex, and is still the subject of research, but it is clear that in some of the Cardington tests the slab acted as a membrane and was supported by the colder perimeter beams and protected columns.
Although composite slabs are normally considered to span in one direction between the steel beams, their behaviour in fire conditions appears to be very different. At ambient temperature, the strength of reinforced concrete slabs is often greater than just its flexural resistance and is enhanced by in-plane load- carrying capacity, which can be utilised after initial flexural yielding has occurred. The first type of in-plane action is compressive membrane action, which depends on the slab having in-plane restraint at its edges. Compressive membrane action occurs immediately after yielding, when deflections are still relatively small; it is very sensitive to the amount of in-plane restraint to the slab. The second form of in-plane action is tensile membrane action, which can occur in slabs with fixed or simply supported edge conditions.
In the case of fixed edge conditions, compressive membrane action occurs first, followed by tensile membrane action when deflections become greater. The tensile forces that develop in the reinforcement need to be anchored at the slab supports. In the case of simply supported edges, the supports will not anchor the tensile forces and a compressive ring beam will form around the edge of the slab. When tensile membrane action develops, failure occurs at large displacements, with fracture of the reinforcement.
Observations of the behaviour of the floor slab in the Cardington tests showed that initially, as the steel beams lost their load-carrying capacity, the composite slab utilised its full bending capacity in spanning between the adjacent cooler members. As displacements increased, the slab acted as a tensile membrane, carrying loads in the reinforcement.
All the structural mechanisms working in the slab rely on the mesh retaining its structural integrity. At some locations in the Cardington building, it was observed that the mesh fractured and there was evidence that it had not been placed properly during construction. Observations following Test 6 showed that, in at least one location, adjacent mats appeared to have been butted up to each other rather than being lapped. In such places, the slab sheared locally but no failure occurred.
In order to mobilise the maximum load resistance of the slab, all the structural mechanisms require the mesh to have a degree of ductility. This allows redistribution of internal forces, resulting in a plastic mode of behaviour. The evidence from Cardington, fire resistance tests and Broadgate is that there is normally sufficient ductility, but the poor mesh placement in one area at Cardington could have initiated a failure. Better site control is required.
In the future, a new type of mesh might be available with much greater ductility. Such a mesh is already used in New Zealand, where it is required for earthquake resistance. Although the Cardington tests and the Broadgate fire have shown that the mesh used in UK appears to be adequate, it is likely that improved ductility would give greater reliability.
5.1.3 Modelling of floor slabs in fire
The Building Research Establishment has developed a simple structural model [7,8] that combines the residual strength of the steel composite beams with the slab strength calculated using a combined yield line and membrane action model. The model has been calibrated against the Cardington fire tests and other test results on slabs. In carrying out this process, some assumptions have been made that are thought to be conservative. The Design Tables presented in Part A are based on the BRE model. It is expected that, in time, the model will be revised to take into account other work being carried out [23,24]. In the future, less conservative design information might be available.
The Design Tables include 150 x 150 mm and 100 x 100 mm square meshes. These meshes have some benefits over the standard 200 mm square mesh when used in composite floors. Cracks at the serviceability limit state are reduced and there is increased health and safety for site personnel casting the concrete (i.e. easier to walk on).
Design points - floor slabs
Floor slabs should be checked as part of floor design zones and their strength should be verified using the design tables in Part A. Alternatively, they can be checked using the BRE method [7,8]. Mesh reinforcement should be lapped properly.
5.2 Steel beams
At both Cardington and Broadgate, the steel beams suffered gross deformation without collapse. Unprotected non-composite and composite beams may reach temperatures in excess of 1000°C in fires. At these temperatures, the bending resistance of the cross section could be reduced by 95% and thus, unless the applied loading is extremely low, the beam effectively ceases to act as a bending element. If collapse is to be avoided, an alternative load transfer mechanism is needed.
Steel beams that are protected to achieve the fire resistance required by building regulations would not generally be expected to reach temperatures at which they would fail in bending. Most beams would be protected so that, in a fire resistance test, their temperature would not exceed about 620°C at the end of the required time period. At this temperature, the beam could be expected to have 60% of its original strength; in all but the most exceptional conditions, it would carry the applied loads. Evidence from actual building fires shows that the beam deformation would probably be small; reuse may sometimes be possible.
In the majority of cases, beams are designed as simple beams and the effects of rotational continuity are ignored. In fire, the behaviour of a beam is affected by its end conditions. Some rotational restraint may be generated but, more importantly, the beam may also be restrained from expanding by the other parts of the structure.
Normally, rotational restraint is beneficial as it causes redistribution of moments, leading to a reduction in the maximum mid-span bending moment occurring in the beam, increasing the fire survival time. However, it was evident from the Cardington fire tests that, due to local buckling near the beam ends, which does not occur in unrestrained standard fire tests, any beneficial effects of rotational restraint cannot be relied upon.
The effects of thermal expansion are unavoidable. A fully restrained piece of steel will yield when heated to temperatures between 100°C and 150°C, depending on the grade. A beam will be subject to both axial and rotational restraint (even if designed to be simply supported). A beam of 9 m span heated to 900°C will try to expand 100 mm. If the temperature distribution across the depth of the beam is non-uniform, it will also bend towards the hotter, lower side. The effect of any axial restraint is to induce compression into the beam. The effect of non-uniform heating, coupled with the rotational restraint, is to induce hogging (negative moments) which also tends to add to the compressive stresses in the lower part of the beam at its supports. These compressive stresses are in addition to those caused by rotational restraint at ambient conditions. The lower part of the web and/or the bottom flange of the beam may yield plastically in compression and then buckle, as observed on some of the beams in the Cardington tests.
Until methods that can predict the effects of buckling and the consequent effect on negative bending resistance are developed, unprotected beams should be treated as simply supported members in any design assessment.
From examination of protected beams at Cardington and from observations of beams following actual building fires, protected beams do not appear to suffer from local buckling and continuity may be assumed.
In the Cardington tests, unprotected beams acting compositely with the floor slab have not been observed to separate from the slab or to slip excessively relative to the slab. From this, it can be deduced that the shear connectors generally perform reasonably well. There was evidence in one of the Cardington tests of the failure of one shear connector on one beam, but it did not lead to a progressive (unzipping) failure of connectors. Generally, the degree of shear connection provided in the composite beams at Cardington was low, leading to the conclusion that degree of shear connection is not a factor that may affect performance in fire. Evidence from other fires confirms this.
The effect of applied load on the deformation of the floors at Cardington has been investigated using finite element techniques. Some studies have shown that the deformation of the structure may be independent of the applied loading . This is because the dominant loads and deformations are caused by thermal expansion. The modelling has shown that the comparatively low loads applied in the Cardington tests may not have contributed significantly to the good performance in the fire tests and that the conclusions can be applied safely to the slightly higher, more usually, assumed levels of design loading.
5.2.1 Edge beams
Edge beams have the additional role of supporting external cladding. Cladding normally falls into one of two types: masonry/brick, which has an appreciable self weight; and curtain walling, which is less heavy. The consequences of damage are different in each case. Another factor that must be considered in assessing consequences of any damage is the number of storeys.
In some tests, the edge beams were unprotected. In these cases, the wind posts acting in tension above the fire compartment were beneficial, resulting in very small vertical displacements of the beams. The design recommendations therefore recognise the role of vertical ties (wind posts) as a means of supporting edge beams.
Masonry cladding is usually supported at each storey from the edge of the floor slab, with the weight being taken by the edge beam. Curtain walling is normally designed to span between floors and is supported from the floor slab at each floor level. With some types of curtain walling, wind posts are built into the mullions to support the wind pressure on the face of the wall. The posts are then connected to the floor slab or the edge beam. This was the case in the Cardington building. It was also the case at Cardington that the wind posts acted as vertical ties, supporting the edge beam and preventing significant vertical deflection, although some twisting of the beams did occur. In the tests, wind posts were shown to be as effective as applied protection in reducing vertical deformation.
During some of the fire tests, the wind posts at Cardington carried a tensile stress of about 50 N/mm2. This stress was low because the posts were designed to resist bending.
Mathematical modelling in which either vertical ties or fire protection to the edge beams has been omitted has demonstrated that the effect on time to structural failure of the floor system is small. The effect of edge beam deformation on the cladding system is thought to be the most important consideration.
All the Cardington natural fire tests (Tests 3, 4, 5 and 6) showed that the temperatures of unprotected steel beams close to openings were lower than those in the body of the compartment.
Visual evidence of a cool `window zone' was provided by Test 5. Although atmosphere temperatures were not measured within 6 m of the windows, it was clear from the appearance of the galvanized steel decking at ceiling level that lower temperatures had been experienced within one metre of the windows. In this zone, on both sides of the compartment, the galvanized decking was hardly discoloured whilst beyond one metre the surface was heavily oxidised. Temperatures of the upper surface of the profiled steel decking taken through the slab at 5 m and 10.5 m from the windows reached levels above the melting point of zinc (420�C) whilst close to the window, at 0.3 m, only 298�C was recorded. This can be seen in Figure B.5.4. In the figure, the lighter colour of the steel decking (towards the left of the figure) shows areas where the galvanizing is unaffected by heat. The effect of the cool window zone on the galvanized steel decking
Lintel beams above windows are subject to asymmetric heating. Heat radiating from surfaces and flames inside the compartment only reaches the interior face of the beam. The exterior face is subject to a far lower level of radiation from the flames that issue from the windows and these flames may also be cooled by cold air being drawn into the compartment. The test data indicate that the temperature rise of lintel beams is about 25% less than that of fully exposed internal beams (see Figure B.5.5).
Figure B.5.5 Lintel beam temperature rises in Cardington test fires were about 25% cooler than those for internal beams
It is not easy to quantify the effect of window openings in reducing temperatures because it depends on the severity of the fire. Conservatively, it may be assumed that lintel beams above window openings are not heated to the same extent as similar internal beams, and a 50°C reduction in temperature may safely be assumed in assessing the thickness of any fire protection or the strength of the beam. However, when using tabulated data for fire protection materials, the beam temperature must be increased by 50°C to obtain a reduction in thickness. This is because the thickness of fire protection given in tables is sufficient to keep the steel temperature below a given temperature for a given time. A reduction in thickness can only be obtained if the table appropriate to a higher temperature is used.
5.2.2 Masonry cladding
Falling masonry is a hazard to firefighters and others in the vicinity of tall buildings. As masonry will be supported from the edge beams or the edge of the floor slab, there is less chance of dislodgement if the deformation at the edge of the building is reduced. Deformation will be restricted if the edge beams are either protected or supported with vertical ties: this will normally occur as a consequence of the design recommendations in Part A.
5.2.3 Curtain walling
Glazed curtain walling does not perform well in fire and will frequently break up (because of heat rather than movement of its support), allowing glass and other debris to fall to the ground. Controlling the deformation of the edge of the slab and edge beam is likely to have little effect on the degradation of most curtain walling systems.
In the rare cases where fire resisting glass is used, its performance should not be compromised by distortion of its support. It is sensible to control deformation of the edge beam by protection or by the use of vertical ties, as for masonry cladding.
Design points - beams
Beams contained within floor design zones may be unprotected.
When designing unprotected beams (in fire), no effect of continuity should be taken into account.
In assessing strength or fire protection thickness, edge beams above windows may be assumed to be 50°C cooler than a similar beam within the compartment.
Unprotected columns, like unprotected beams, may reach temperatures in excess of 900°C in fires. At these temperatures, the buckling resistance will be reduced to virtually zero, so any support to the floors above will be lost.
Although local squashing of columns was observed at Cardington and in the Broadgate fire, no collapse occurred because the structure had the ability to carry the column loads using alternative load paths. It is very likely that the same would be the case in most composite steel-framed buildings. The exception is when the column is a `key' element and essential to the stability of the building.
Although none of the protected columns in the Cardington tests failed, analysis of the strain gauge measurements showed that there were large moments in some of the columns after the fires. These moments were caused by two additive effects, expansion of the heated floor relative to other floors and induced P-delta effects.
As a fire-affected floor structure heats up, it expands. Normally the structure is anchored horizontally at the core areas, so the movement tends to be greatest away from cores and at the edges of the floor. Horizontal displacements of the floors above and below the fire-affected floor are much smaller, so columns between these floors are displaced horizontally in the middle of a two-storey length thus inducing bending moments. In addition to the moments induced by floor movement, moments will be induced by the P-delta effect as the vertical load carried by the column is displaced. Analysis carried out by BRE  has shown that, in some circumstances, the failure temperature of an affected column might be reduced and additional protection would be required.
Columns in multi-storey buildings are generally not heavily loaded. The load ratio, as defined in BS5950-8, rarely exceeds 0.5. The effect identified by BRE increases the applied loading and load ratio and thus reduces the limiting temperature. To allow for this effect, it is recommended that the fire protection should be based on a limiting steel temperature of 500°C rather than 550°C or more. A reduction to 500°C is equivalent to increasing the steel stiffness and strength by at least 25%.
Most column fire protection is based on limiting the steel temperature to 550°C. However, in many cases, the minimum practical thickness of protection that may be applied to columns is sufficient to achieve 60 minutes on columns with a section factor not greater than 200 m-1. As a result of this, for 60 minutes, the thickness will normally be sufficient to keep the column temperature well below 500°C. For 30 minutes fire resistance, temperatures will often be well below 400°C.
In one of the Cardington tests, the vulnerability of partially protected columns was demonstrated when the unprotected top part of two columns squashed. No collapse occurred but the floors above the fire compartment were all affected. It was concluded that, until more was understood about how much of a column could be left exposed, in the subsequent tests and in any design recommendations, columns should be protected for their full height. Protecting columns over their full height is important if damage is to be confined to the fire floor.
Design points - columns
For 30 minutes fire resistance, columns should be protected up to th e underside of the deepest supported beam.
For 60 minutes fire resistance, columns should be protected up to the underside of the floor slab.
The fire protection to a column should be based on a reduced limiting steel temperature. The amount of protection should be based on a temperature of either 500°C or 80°C less than that derived from BS5950-8, whichever is the lower.
Although connections in braced multi-storey frames are normally designed to transmit shear forces only, they do in fact have a significant moment capacity that is not utilised in ambient temperature design. The rotational stiffness of these `simple' connections can be significant. In fire, the connections have the potential to develop moments that could enhance the load-carrying capability of a nominally simply supported beam. However, in the case of the Cardington tests, it was found that local buckling of the unprotected beams occurred adjacent to the connections in many instances, which negated the ability to develop negative moments. Further investigation of the effect of local buckling on the moment capacity of connections will be required before redistribution of moments in fire-affected steel beams could be relied upon.
During the heating phase of the tests, connections performed well, however some connections suffered partial failure during the cooling phase. This was because the beams are subject to large compressive stresses during the heating phase; these arise from restrained thermal expansion and thermally induced curvature. The restraints had the effect of causing plastic compressive deformation and, in some cases, local compressive instability. Both of these result in shortening of the beams. As the beam cools, the plastic strains cannot be recovered and the beams are therefore shorter than before the fire. The overall effect is similar to the `lack-of-fit' problems with which engineers are familiar. The connections were thus subject to tensile forces, which in some instances led to some form of failure.
Three types of simple connection, end-plate, fin-plate and cleated are considered in detail below.
5.4.1 End plate connections
Observations of the performance of end-plate connections in the Cardington tests have shown that these connections perform well in fire. There was no complete failure in any connection involving end plates. However, during the cooling phase following the fire, a number of effects were observed. These included rupture of one side of the end plate close to the heat-affected zone of the end plate to beam web weld and, in one case, failure of the web as the end plate was torn off the end of the beam. Both types of failure were due to high tensile forces generated by contraction of the beams during cooling.
In the case where the plate was torn off the end of the beam, no collapse occurred because the floor slab transferred the shear to other load paths. This highlights the important role of the composite floor slab with proper lapping of the reinforcement.
In the case of the plate rupturing, only one side of the plate became detached leaving half the original number of bolts in place to resist shear (see Figure B.5.6). The floor slab may also have distributed some shear to other load paths in this case.
Figure B.5.6 Schematic of end plate rupture
5.4.2 Fin-plate connections Fin-plate connections were used in the Cardington tests to connect the secondary beams to the primary beams. These connections appeared to perform well in fire but some problems were encountered on cooling. It was found that the tensile forces generated in the beam due to plastic strain at high temperatures sheared the bolts in the fin-plate connections at one end of the beam. No collapse occurred, and the shear was carried by a combination of the secondary beam resting on the bottom flange of the primary beam and the shear resistance of the composite slab. Local buckling in the bottom flange of the beam was also experienced in this type of connection, as shown in Figure B.5.6.
Tensile tests at ambient temperature show that fin-plate connections have a large amount of ductility. A major contribution is the elongation of the bolt holes. At Cardington, the bolts fractured, with no apparent elongation of the bolt holes (see Figure B.3.7), indicating that the bolt material had lost strength during the fire. Although bolt temperatures were not measured, an unprotected fin-plate in Test 6 reached 862�C.
Fin-plate connections may be detailed in two ways. At Cardington, they were used for beam-to-beam connections and in all cases the top flange of the incoming beam was notched back. Fin-plate connections are also commonly used for beam-to-column connections, especially when the column is a structural hollow section. In this case, it will often be unnecessary to notch the top flange of the beam and, should the bolts fail, the vertical displacement of the beam will be limited because the underside of the top flange will bear on the top edge of the fin plate. In combination with a composite floor slab, this is a safe mode of failure that will permit longitudinal movements of the beam while maintaining vertical support.
Although connections formed from notched beams will not have the same fail- safe mode of behaviour as those utilising unnotched beams, they are nevertheless considered adequately safe because there is the possibility of an incoming beam resting on the bottom flange of a main beam, and the adoption of the slab design model (see 5.1.3) will often result in larger reinforcing mesh than was used at Cardington.
5.4.3 Cleated connections
Although there were no cleated connections used in the Cardington frame, SCI has conducted a number of tests on composite and non-composite cleated connections in fire . These connections consisted of two steel angles bolted to either side of the beam web using two bolts in each angle leg, then attached to the flange of the column also using two bolts. The connections were found to be rotationally ductile under fire conditions and large rotations occurred. This ductility was due to plastic hinges that formed in the leg of the angle adjacent to the column face. No failure of bolts occurred during the fire test. The composite cleated connection had a better performance in fire than the non- composite connection.
It is likely that the ability of the angle cleats to deform will provide the necessary ductility on cooling.
5.4.4 Bolts and welds
Observations of the connection behaviour at Cardington indicate that bolts or welds do not fail prematurely in fire situations and it is not generally considered necessary to design welds or bolts specifically for fire scenarios. However, the material properties of the bolts and welds are known to vary in a similar way as those of structural steel . Current guidance on the strength reduction for bolts and welds given in BS 5950-8 allows 80% of the strength reduction factor used for structural steel (at 0.5% strain). No guidance is given in the Eurocodes. In the Cardington tests no failures of fasteners occurred during the heating phase of the fire. However, during the cooling phase, a number of bolts acting in single shear in fin-plate connections failed due to the tensile force generated by thermal contraction as described above.
Design points � connections
At the ends of unprotected beams, connections with the ability to deform as the beam shortens on cooling are preferable to less deformable types.
5.5 Overall building stability and bracing
The complete collapse of a multi-storey building as a result of a fire is never acceptable, so consideration must be given to overall building stability. The design recommendations are concerned principally with the performance of non-sway frames, which would usually be braced by a steel bracing system or by some form of brick or reinforced concrete shear walls. The latter two forms of shear wall can be expected to have a high degree of inherent fire resistance. Bracing systems formed from steel have been shown to perform well in fires but their design and location needs some consideration.
The Cardington building had three braced cores. All bracing members were positioned in protected shafts and were thus shielded from any of the test fires. It is normal to place bracing systems inside shafts to isolate them from potential fires, in which case it is not necessary to protect the individual members.
The cost of fire protecting bracing members is often high because the members are comparatively light and therefore have high values of section factor (A/V). Bracing within a structure has two roles. It resists lateral wind forces but, especially for tall buildings, it contributes to the overall stability of the structure. In fire, it is important to recognise these two roles. In the case of wind, some guidance is offered by the structural design codes [3,10]. BS 5950-8 recognises that it is highly unlikely that a fire will occur at the same time as the building is subject to the maximum design wind load; it consequently recommends that, for buildings over 8 m tall, only one-third of the design wind load need be considered and, for buildings up 8 m tall, wind loading may be ignored. None of the codes gives guidance on overall frame stability, but it is self evident that the designer must consider it for multi-storey buildings.
The design recommendations in this guide are based on the assumption that sway instability will not occur. Bracing will normally be fire protected but there can be justification for reducing the amount of fire protection on bracing by taking into consideration the following: * Bracing may be shielded from fire by installing it in shafts or within walls. The shielding should provide all the necessary fire protection. * Masonry walls, although often designed as non-load-bearing, may provide appreciable shear resistance in fire. Bracing systems are often duplicated and loss of one system may be acceptable.
Design points � bracing
Whenever possible, bracing should be positioned in protected shafts or built into walls, to avoid having to protect individual members.
5.6.1 Structural considerations
One of the key strategies of fire safety is to contain the fire in its compartment of origin. In most multi-storey buildings, each floor forms a single compartment; within a floor, the walls are not generally compartment walls. However, in some buildings, the design does assume that certain walls that subdivide a floor are compartment walls.
From the observations at Cardington, it is clear that large displacements of the floor above a wall can occur and this could affect the performance of some walls. It is common practice to use lightweight walls that are constructed as stud walls using fire-resisting boards fixed to cold formed channels. These walls are not designed to be load-bearing and have little resistance to the potential loads that could be imposed from a deflecting structure during a fire. Blockwork walls are less commonly used but they are much more robust than stud walls and could be expected to resist quite large loads, even if designed as non-load-bearing. However, the presence of thermal gradients through masonry walls may cause out-of-plane bowing towards the heat source where the wall is simply supported top and bottom, and away from the heat source where the wall cantilevers from the base. Out-of-plane thermal bowing may also occur. The material properties of the masonry will also deteriorate at high temperatures, which may cause reverse bowing as a result of eccentricity of the load. Because of the precautions taken in the Cardington tests, no failures or near failures occurred in the masonry compartment walls.
For Tests 4 and 5 on the Cardington frame, internal compartment walls were constructed using steel-framed stud partitions with fire resistant board. In Test 4, the walls were placed on the column grid under unprotected beams. Due to the shielding effect of the wall, the vertical deflection of these beams was very small and the integrity of the compartment wall was maintained. In Test 5, the compartment wall was placed `off-grid' and under the soffit of the floor slab. The deflection of the slab caused integrity failure of the compartment wall.
It is advisable to place compartment walls under beams whenever possible. However, to comply with the insulation criterion for separating elements as specified in BS 476, these beams would normally be protected, thus incorporating an additional level of safety in maintaining the integrity of the compartment wall.
Walls positioned off column grid lines require careful detailing to accommodate large vertical displacements. The deflection of a beam may be of the order of span/30. In the design recommendations, it is presumed that this deflection exists over the central 50% of the span, with a linear reduction from this value to zero at the supports. This is a much more onerous requirement than current practice.
The measurements in Test 4 indicate that the temperature of one side of an exposed beam above a wall surrounding the fire compartment is significantly lower than that of a fully exposed beam in the centre of the compartment. The data indicates that a beam above a wall on the compartment perimeter shows a similar temperature reduction to that observed in a lintel beam above a window (Section 5.2.1). In addition, the fact that the beam is only exposed on one side and perhaps subject to less convection in corner `dead zones' reduces the temperature rise still further. Measured temperatures in Test 4 (see Figure B.5.5) showed that this additional reduction was about 20%, between 100 and 200�C in this case.
Beams above compartment walls (not partitions) require protection to meet the insulation criteria, as the surface temperature on the unexposed side is likely to exceed the allowable limits for insulation specified in BS 476.
Design Points � Compartmentation
Whenever possible, compartment walls should be positioned on column gridlines.
Compartment walls that are crossed by an unprotected beam should be designed to accommodate the possible vertical displacement of the structure during a fire.
5.7 Other influencing factors
In this Section, three factors that influence the behaviour of steel-framed buildings in fire are discussed: robustness, sprinklers and suspended ceilings.
Steel-framed multi-storey buildings are designed to meet certain robustness requirements. Although this design guidance does not make reference to it, the excellent robustness of steel-framed multi-storey buildings underpins the design philosophy.
The good performance of the structure in the Cardington tests may be directly related to the inherent robustness of composite steel-framed buildings. In the UK, buildings over five storeys are required to be able to resist disproportionate collapse (for example, from gas explosions). Smaller composite buildings, although not required to have the same degree of robustness, will also be quite robust. Robustness exists through the continuity of the floor slab and connections.
The expected performance (or robustness) of structures in fires can be seen as more onerous than the robustness requirements to resist explosions. In a multi- storey building fire, a structure is expected to resist the effects of fire without collapse and to contain the fire within the compartment of origin. The area of structure that might be engulfed in the fire may often be larger than that required to be considered for an explosion. A fire is considered to `fill', the compartment, which could be a complete floor. The area allowed to be affected by local collapse and debris loading in an explosion is 210 m2, provided that progressive further collapse does not occur. Each floor at Cardington was 945 m2; most buildings will have fire compartment areas larger than 210 m2.
The performance expected of buildings in fire is therefore onerous compared with some other accidental design conditions; nevertheless, buildings designed using the proposed recommendations will meet the higher safety standards expected in fire.
`Active fire protection systems' is a general term used to describe systems designed to control or extinguish fires or to alert people to the presence of fire. The system that makes the greatest contribution to structural fire safety and property protection is an automatic water sprinkler system.
The design guidance does not rely on the presence of sprinklers but the installation of sprinklers will have a beneficial effect on performance. The value of sprinklers is well known and many buildings will have sprinklers installed.
Traditionally, passive and active fire protection measures have been considered as independent components of an approach to fire safety. The use of active protection systems has been promoted by insurers, who recognise the value of such systems in controlling fire growth and therefore reducing the extent of fire damage in buildings. Statistical information from insurance sources confirms that sprinkler systems can play an important role in reducing financial loss.
5.7.3 Suspended ceilings
It is not well known that certain types of non-fire-rated ceilings can shield the floor and beams from the extremes of the fire, and will therefore contribute to the performance of the structure.
The natural fire tests undertaken by BHP Australia (see Section 4.3) showed clearly that even non-fire-rated suspended ceilings can significantly reduce the temperature to which beams are exposed (see Figure B.5.7). In both the William Street and Collins Street tests, cast plaster tiles were used, and in the latter case the tiles were described as having a fibreglass backing blanket. In both tests, the tiles were supported on steel runners rather than aluminium extrusions (which would melt at 660�C). Most significantly, it demonstrated that even though the ceiling was breached, its presence was sufficient to reduce the atmosphere temperature in the region of the beams by around 400�C. The unprotected beams reached temperatures between 280�C and 470�C, measured at the lower flange position; in similar fire conditions at Cardington, but without the ceiling, temperatures over 1000�C were recorded.
The reason for this difference becomes apparent from the time-temperature plots (see Figure B.5.8,Figure B.5.9 and Figure B.5.10), which are reproduced from the BHP report. They show that the ceiling began to lose its integrity at around 30 minutes, just before the fire temperature peaked at 33 minutes, and that beam temperatures did not begin to rise until after the fire temperature had passed its peak. The beams were thus only fully exposed to a declining fire, not the growing fire that the standard BS 476 fire simulates.
Although it is difficult to quantify the effect of non-fire-rated ceilings with steel runners, it is clear that they could be expected to reduce beam temperatures and thus enhance stability significantly. The ceilings, in combination with other fire safety provisions, could justify relaxation of structural fire resistance requirements.
Figure B.5.7 Suspended ceiling before and after the Collins Street test (BHP, Australia)
Figure B.5.8 Maximum atmosphere temperatures in Collins Street test (BHP, Australia)
Figure B.5.9 The effect of a suspended ceiling on atmosphere temperatures (BHP)
Figure B.5.10 The effect of a suspended ceiling on steel beam temperatures (BHP, Australia)
6 THE WAY FORWARD
In this publication, design recommendations are presented that will allow many secondary beams in composite steel-framed buildings to be designed without applied fire protection. The recommendations can be seen as Level 1 recommendations; they are simple to apply and are conservative.
The design recommendations described are based on the accepted concept of fire resistance given in building regulations. As knowledge of real fire behaviour develops and research leads to a better understanding and quantification of the risks and consequences of our design decisions, it should be possible to design buildings to resist realistic fires and not for notional periods of fire resistance.
It is hoped that, in time, it will be possible to offer all consulting engineers and architects more refined fire engineering design aids (Level 2) to enable them to take full advantage of the real behaviour of composite steel-framed buildings in fire.